Used Double Track Bombardier in Ny State

Railway bridges

A. Pipinato , R. Patton , in Innovative Bridge Design Handbook, 2016

4.7 Construction process

Railway bridges are specific structures requiring a deep knowledge not only of structural engineering, but also of the operations and safety requirements of lines used daily by great numbers of passengers. For this reason, the following considerations arise:

•

Timing: When dealing with the construction or replacement process in railway engineering, the driving factor of the design is track time: so operations, maintenance, and new construction are all relevant factors that should be carefully evaluated by the designer in order to produce the best bridge in the shortest time. For this reason, the design efficiency could be sacrificed for a shorter construction period.

•

Simplifying construction: Simple constructions are generally preferred to complex solutions, and elements such as simple spans and bolted construction (i.e., stiffeners bolted to web plates) are still widely used for railway bridges, whereas continuous spans with welded stiffeners are standard practice in highway bridge design. The use of bolted construction reduces fatigue requirements, and simple spans allow the replacement of each individual span, thus minimizing traffic interruptions.

•

Precasting: A direct consequence of the previous points is that it is becoming a common practice to design and erect spans in nearly complete form in order to expedite span realization. Steel spans and precast concrete box beams, as well as other superstructure types, may be shipped to a construction site fully assembled in order to lift all in place quickly and restore traffic as soon as possible (AREMA, 2013).

•

Material savings: Often the outcome of a design is counterintuitive to the standard practice of producing highly efficient structural systems that use a minimum amount of material. In the long term, this break from common practice proves more beneficial to railway companies due to the savings yielded from a design that lasts many years, requires minimal maintenance, and provides a construction period that keeps trains moving (AREMA, 2013).

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Research and Training in ABC Structural Systems

Mohiuddin Ali Khan Ph.D., M.Phil., DIC, P.E. , in Accelerated Bridge Construction, 2015

3.8 ABC for railway bridges

According to the University Transportation Research Committee (UTRC) at Rutgers University, the population and employment in the Northeast is expected to grow some 35% over the next 25   years, bringing the reality of even more traffic and congestion. It is the problem of mega cities everywhere.

For example, the Washington–Boston Northeast Corridor rail line does not have the ability to absorb the crush of new travelers in the coming decades. It will require a significant investment to grow capacity, improve reliability, and serve new markets. The Federal Railroad Administration (FRA), (an agency within the U.S. Department of Transportation) is leading the development of a Passenger Rail Corridor Investment Plan (NEC FUTURE) to define the investment required to keep the Northeast Corridor vibrant and to prepare the roadmap for federal, state, and private investment. Rebecca Reyes-Alicea (FRA's program manager for NEC FUTURE), is leading the effort for the planning process, the challenges in working on a corridor that crosses eight states and is served by commuter, intercity and freight railroads, and the steps NEC FUTURE is taking to develop a long-range investment program.

To expedite the construction of railroad bridges in a congested area, it is expected that ABC procedures will be applied. The requirements of repair and rehabilitation methods for railway bridges are based on the AREMA (American Railway Engineering and Maintenance-of-Way Association) Code. D-B management methods and prefabricated construction are applicable. Coordination with railway and transit authorities will be required for catenary construction and signaling, etc., to ensure the rapid availability of train service.

Infrastructure upgrades are central to ensuring safe and reliable performance, including the rehabilitation, expansion, and replacement of the bridges. The infrastructure upgrade program will also provide for track renewal, passenger communication upgrades, signal system upgrades, and improvements to overhead power lines and electric substations. Guidelines for rating railway bridges are given in the AREMA Manual. A variation of Cooper E loading is used. Maximum design live loads for replacement bridges are Cooper E80, but the bridges need to be rated for Silver liner, Bombardier, and other types of locomotives being used by the transit agencies. Unlike FHWA design methods, only the conservative allowable stress method is used. Inventory and inspection forms are different from the SI&A Sheet.

Railway bridges have several differences as compared to highway bridges, and are fewer in number:

•

Deck slabs are not present since train wheels are supported on rails, sleepers, and ballast.

•

Steel bridges are generally of the through-girder type with floor beams and require lateral bracing. Timber bridges are supported on timber-framed trestles or pile trestles.

•

Stone arch and masonry bridges of smaller spans than highway bridges are common.

3.8.1 Storm-resilient grid for New Jersey transit system

(Reference Reuters Report by Selam Gebrekidan and Leslie Gevirtz).

New Jersey sustained a severe blow when Sandy made landfall. The storm also cost New Jersey Transit an estimated $400   million.

The Department of Energy and the state of New Jersey plan to design a small electric grid that will serve the state's transit system and withstand the onslaught of storms like Superstorm Sandy.

The microgrid will power the transit system's rail operations between Newark, Jersey City, and Hoboken in New Jersey. It will be designed by the energy department's Sandia National Laboratories, which has worked on such grids for the U.S. military.

The project will make a key part of the northeast energy infrastructure resilient to changes brought on by climate change.

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Seismic Bridge Design

Mohiuddin Ali Khan Ph.D., P.E., C. Eng., M.I.C.E. (London) , in Earthquake-Resistant Structures, 2013

8.9 Comparison of Highway and Railway Bridges

Railway bridges have higher risks due to concentration of passengers and are more conservatively designed than highway bridges and have historically performed well in seismic events with little or no damage for the following reasons:

•

The track structure functions very effectively as a restraint against longitudinal and transverse movement.

•

The spans are small.

•

Heavy concrete decks are not present and dead load inertia forces are smaller.

•

The types of damage that are permissible are very limited compared to highway bridges.

•

Post-seismic event operation guidelines put restrictions on train traffic and speeds depending on the intensity of the event, until proper inspection is carried out.

•

Level 1 ground motion has a smaller return period for railway bridges: 100 years as compared to 450 years for highway bridges. Acceleration coefficients are expressed as a percent of gravity for return periods of 50, 100, 250, 475, and 2,400 years.

•

Risk factor is an integral part of seismic design for railway bridges. For level 2 and 3 ground motions, risk criteria are based on a high passenger train occupancy rate.

•

The seismic response coefficient (Cm) is multiplied by a damping adjustment factor usually greater than 1, resulting in higher seismic force.

•

Analysis is based on serviceability and ultimate and survivability limit states, unlike LRFD analysis for highway bridges.

•

The substructure response modification factor (R) is smaller for level 3 ground motion, resulting in higher seismic design moments.

•

Train live load is combined with dead load to give higher horizontal forces.

A number of bridges located in New Jersey, New York, and Pennsylvania were analyzed by the author. The findings with regard to analysis procedures have contributed to a chapter, "Seismic Analysis and Design" in the proposed New Jersey LRFD Bridge Design Manual. By way of illustration one study is discussed in the following section. All four bridges have acceleration factors of 0.18.

8.9.1 Case Study: Route 46, New Jersey

Figure 8.8 shows a plan of U.S. Route 46 over Peckman's River in New Jersey. The bridge replacement was required due to scour from Hurricane Floyd. In New Jersey, all counties are classified as Zone 2. Hence, a quasi-static method rather than a dynamic analysis, and such response spectrum is not required.

Figure 8.8. Case study of an integral-abutment bridge in New Jersey designed by the author.

The uniform load method is applicable to regular bridges with acceleration coefficients of 0.10 and 0.18. The method is described in AASHTO (AASHTO 1996, 2). The procedure used is described in detail in a paper by Khan (2004).

•

From single-mode coefficients, static displacement Vs=ρ L/k; γ=βVs

•

Compute the time period T=2π(γ/ρ.α.g)^0.5

•

Elastic seismic response coefficient Cs=(1.2AS)/(T)^2/3 where A=acceleration coefficient, S=site coefficient, and T=period of oscillation.

A limiting value of 2.5 for A was used.

•

For single-mode analysis, using deck weight Vs and single-mode coefficients (α),(β),(γ) factors

•

An equivalent static earthquake loading pe(x)=β(Cs)(w(x) Vs(x)/γ was computed and modified by the response modification factor R.

•

Seismic force=pe(x)L/R.

Seismic load combinations used:

•

(DL+100% longitudinal+30% transverse)

•

(DL+30% longitudinal+100% transverse)

Unlike conventional single-span bridges where seismic forces are neglected, there is a need to consider seismic behavior in single-span bridges with integral abutments to aid in the design of the buried abutment piles.

Freedom of top of pile movement for thermal and seismic loads is provided by placing the top unsupported 5   m length in a steel pipe. Only a single row of piles is used. For pile design, seismic moments and forces are combined with dead load and partial live load results. Since the pile minor axis is placed perpendicular to the direction of beam, it is important to check the total unsupported pile length of pile, for axial buckling.

For bridge stability, efficient details of integral connections between pile cap (with single row of piles) and beam ends are required. Connection details from various state bridge design codes were studied and compared.

The following observations resulting from the preceding study and the others can be made:

•

Seismic performance is typically better in systems with regular configurations than in those with skew, unequal pier heights, and curved bridges.

•

The equivalent static analysis method is suitable only for preliminary design. Final design moments and forces should be checked by spectral methods. AASHTO's LRFD code specifies R Factors, which are rounded off, that appear to be approximate. Since this factor scales down the design moments, future research should focus on arriving at a more accurate assessment of the R Factors.

•

An independent code committee should verify the suitability/accuracy of each standard software program used in this type of analysis.

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Rapid Bridge Insertions Following Failures

Mohiuddin Ali Khan Ph.D., M.Phil., DIC, P.E. , in Accelerated Bridge Construction, 2015

6.1.2 Maintenance of old bridges

We are used to maintaining our cars on a regular basis, replacing essential parts every few years and replacing an old car every 10–15   years. The same approach applies to bridges. We replace the concrete or asphalt topping or the deck slab itself every 10–15   years. Most bridge superstructures are replaced every 50–75   years with or without the substructure.

Bridges are subjected to repeated wear and tear from heavier trucks and from extreme events such as flood and earthquakes. Maintenance requires timely rehabilitation, repair, and retrofit. There is an old saying that "a stitch in time saves nine." If a bridge is neglected, an emergency replacement may result, and that is where ABC is particularly useful.

Life cycle costs are likely to be higher than the initial investment. It may be more economical and easier to design a new bridge than to maintain the same bridge over its remaining life.

Not counting railway and transit bridges, the three types of roads on which bridges are located are:

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Interstate: Due to heavy average daily traffic (ADT), not even a single lane can be shut down for maintenance. Already, there are traffic jams during rush hour. Time loss is a colossal waste at national scale.

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Collector: A lane can be closed for a short duration with nighttime work using ABC.

•

Local: When ADT is low, ABC is not essential, and for small spans modular bridges can be used.

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Applied Loads and Stability of Steel and Steel-Concrete Composite Bridges

Ehab Ellobody , in Finite Element Analysis and Design of Steel and Steel-Concrete Composite Bridges, 2014

3.4.3.1 Braking and Acceleration Forces

Similar to railway bridges, highway steel and steel-concrete composite bridges are subjected to braking and acceleration forces. According to EC1 [ 3.1], braking force, Q _lk, shall be taken as a longitudinal force acting at the surfacing level of the carriageway. The characteristic value of Q _lk limited to 900   kN for the total width of the bridge and should be calculated as a fraction of the total maximum vertical loads corresponding to the Load Model 1 likely to be applied on Lane Number 1, as follows:

(3.16) $\begin{array}{l}{Q}_{1\mathrm{k}}=0.6{\alpha }_{\mathrm{Q}1}\left(2Q_{1\mathrm{k}}\right)+0.1{\alpha }_{\mathrm{q}1}{q}_{1\mathrm{k}}{w}_{1}L\hfill \\ 180{\alpha }_{\mathrm{Q}1}\left(\mathrm{kN}\right)\le Q_{1\mathrm{k}}\le 900\left(\mathrm{kN}\right)\hfill \end{array}$

where L is the length of the deck or part of it under consideration. For example, Q _lk  =   360   +   2.7L (≤   900   kN) for a 3   m wide lane and for a loaded length L  >   1.2   m, if α factors are equal to unity. The upper limit (900   kN) may be adjusted in the National Annex. The value 900   kN is normally intended to cover the maximum braking force expected to pass over the bridge. Horizontal forces associated with Load Model 3 should be defined where appropriate. This force should be taken into account as located along the axis of any lane. However, if the eccentricity effects are not significant, the force may be considered to be applied only along the carriageway axis and uniformly distributed over the loaded length. Acceleration forces should be taken into account with the same magnitude as braking forces, but in the opposite direction.

In the United States, AASHTO [1.23,1.24] specifies that the braking force on highway bridges shall be taken as the greater than 25% of the axle weights of the design truck or design tandem or 5% of the design truck plus lane load or 5% of the design tandem plus lane load. The braking force shall be placed in all design lanes carrying traffic heading in the same direction. The forces shall act horizontally at a distance of 1.8   m above the roadway surface in either longitudinal direction to cause extreme force effects.

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Fatigue in railway bridges

M . Gilbert , in Fatigue in Railway Infrastructure, 2009

3.5 Metal and concrete bridges

3.5.1 Metal bridges

Clearly, existing metal railway bridges take a diverse range of forms, ranging from the impressive or historically important (e.g. Brunel's Saltash Bridge, the Forth Rail Bridge, Stephenson's remaining tubular bridge at Conway, etc.) through to the numerous workmanlike short- and medium-span beam bridges distributed all over the network.

To give an indication of the numbers of metal bridges of different types on the network, of a sample of 2236 overline rail bridges assessed as part of the 'BridgeGuard' programme between 1995 and 2000, 16% were metal girders (including half through girders), 5% were of jack arch construction, 2% comprised trough decks, 1% cast iron beams and 0.4% were metal trusses; the majority of the remaining bridges assessed were masonry arches, with some concrete bridges also.

In the case of the numerous beam bridges on the network, these could be designed using scientifically based methods from when they were first constructed (though fatigue was initially not well understood). Stresses in beams subjected to given working loads could be calculated and checked against suitably factored maximum permissible values.

For bridges which pushed the boundaries, additional studies were sometimes required. For example, in the case of Robert Stephenson's large tubular rail bridges at Conway and the Menai Straits, prototype models were built and proving tests performed. ³⁰ These identified thin plate buckling in the compression zones of the tubular section as a failure mode (very different from the tensile zone failures which had been encountered previously, when cast iron beams were tested). Thus suitable stiffeners were added to reinforce the compression zone.

In the case of more conventional beam bridges, the main problems which arose were often due to poor detailing (e.g. bridges often had inadequate and/or unmaintainable connections between the main beams and the deck).

By the second half of the twentieth century welding had superseded riveting as the most common connection method. This period also saw the introduction across the network of a number of standard underline bridge designs for short and medium spans (initially approximately 6   m to 30   m). The original standard designs used steel plate girder main beams with steel cross-beams embedded in reinforced concrete floor slabs (e.g. Fig. 3.15). Since the initial standard designs were produced, other standard solutions have also been developed (e.g. incorporating box girder beam sections).

3.15. Typical standard underline rail bridge design (bridge type 'A').

Clearly, standardisation reduced initial design and construction costs and can now potentially reduce assessment costs, since the need for detailed investigations merely to identify construction details on a bridge-by-bridge basis should be obviated.

However, unfortunately most metal bridges on the UK network predate standardisation, with the majority dating from between 1880 and 1920; ¹ many are considerably older. Assessment of these very old structures can be quite difficult. For example, many existing overline spans on the railways comprise jack arches spanning between longitudinal metal beams, but the structural performance of this system is still not particularly well understood. The same is also true for lattice girder station footbridges, which appear to have performed better in practice than is suggested by simple structural analysis.

Despite gaps in our knowledge and understanding of old structures, fortunately failures are comparatively rare. However, this must not lead to complacency, and also it is of paramount importance that when failures inevitably do occur, these should be carefully analysed and lessons learnt. For example, derailment of freight wagons led to a spectacular failure in 2003 of an 1852 bridge at Cahir in the Republic of Ireland. Here derailed laden freight wagons impacted on and subsequently dislodged transverse beams over a large part of the bridge span (Fig. 3.16). While it was initially speculated that the derailment may have been caused simply by spreading of timber way beams which had become rotten (way beams span longitudinally over the transverse beams and should provide continuous support for the rails), an inquiry identified a more complex cause: dynamic interaction between the particular rigid wagons being pulled at the time, travelling at 40 mph, and the variable-stiffness bridge beams, which allowed a wagon to jump the tracks. A specific outcome has been that the speed limit for the bridge (now repaired at a cost of €2.6m) has been reduced when these wagons are being pulled. ³¹ More generally, it is hoped that awareness of the issue of dynamic interaction has increased.

3.16. Cahir Viaduct (1852), Ireland, which partially collapsed while carrying a freight train, 2003 (photo: Aidan Brosnan).

3.5.2 Concrete bridges

Of the concrete overline bridges assessed as part of the 'BridgeGuard' programme between 1995 to 2000, about half were of reinforced concrete, the remaining half being of pre-stressed concrete (together making up 13% of all bridges in the sample). Reinforced concrete was heralded as a largely maintenance-free material when first introduced. However, there have been durability problems, as will be mentioned in Section 3.5.4. Precast, pre-stressed, concrete bridge decks became available after the Second World War. Because these were precast in factory rather than site conditions, the quality of the resulting construction was often excellent (though some early beams did have inadequate shear reinforcement). Post-tensioned beams have been more problematic, primarily due to problems in grouting the tendons on site (any voids left have allowed corrosion).

3.5.3 Fatigue in metal bridges

The fact that subjecting a piece of metal to many cycles can produce failure at a much lower load than would be required for static failure has been recognised by engineers since the mid-nineteenth century. At that time, as evidence of the phenomenon grew, the Board of Trade in the UK imposed more stringent limits on maximum permissible stresses (e.g. 5 tons/in ² for wrought iron). ³⁰

Obviously, understanding of metal fatigue has grown very significantly since then. It is now known that the fatigue process is the result of several effects operating in sequence: initiation of a microscopic defect, slow incremental crack propagation and final unstable fracture.

In practical terms, modern bridge designers normally try to design out fatigue by ensuring that the stress range in each loading cycle is not too large. Additionally, welded connections are particularly prone to fatigue failures, and weld details need to be carefully specified. However, many early metallic bridges have design details which do not meet current fatigue criteria, and the severe consequences of failure mean that these continue to represent a risk.

3.5.4 Corrosion

In addition to fatigue, corrosion is the other main problem affecting metallic bridges. Of the materials used on the rail network, cast iron is the most resistant to corrosion, followed by wrought iron and then steel, the least resistant material. Certain areas of bridges are particularly vulnerable, for example the bottom flanges of plate girders or at footway level in the case of overline bridges with large plate girder edge beams.

In general, corrosion can be prevented by using an effective paint system. Though there will be situations when it is not cost-effective to paint a bridge if the latter is known to have a limited remaining lifespan, unfortunately rusting bridges look unsightly and this strategy tends not to be appreciated by the general public!

In the case of concrete bridges, corrosion is also a significant problem. In early concrete bridges the reinforcement was often not adequately protected by concrete cover. Additionally, the composition of the concrete was often undesirable (e.g. high water : cement ratio, use of unwashed marine aggregates, etc.). De-icing salts applied in cold weather to the road surface of overline bridges are a particularly common cause of corrosion in steel reinforcement.

Many techniques have been developed for repairing concrete bridges but unfortunately most are labour intensive, time-consuming and hence relatively expensive.

3.5.5 Assessment of metal and concrete bridges

In 2001 a new limit-state assessment code of practice was introduced for UK railway bridge assessment, ³² superseding the permissible stress approaches used since the emergence of the railways (except in the case of cast iron bridges). Limitstate approaches are already commonly used in the design of new structures; these properly distinguish between ultimate (collapse) and serviceability conditions (i.e. 'limit states'), and rationally allocate partial safety factors according to the degree of uncertainty associated with each individual parameter (e.g. dead load, material strength, etc.). This represents a very significant step in the modernisation of the assessment process for UK rail bridges, albeit at the expense of slightly increased complexity and hence burden on assessment engineers.

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The Structure of the Product Design Process

Anil Mital , ... Aashi Mital , in Product Development (Second Edition), 2014

3.3.3 The scale effect

In the late 1800s, a railway bridge across Scotland's Firth of Tay swayed and collapsed in the wind. Seventy-five passengers and crew on a passing night train died in the crash. It was the worst bridge disaster in history. So, when engineers proposed bridging the even wider Firth of Forth, the Scottish public demanded a structure that looked like it could never fall down. Chief engineers Sir John Fowler and Benjamin Baker came up with the perfect structural solution: a cantilevered bridge, with a span of 8276  ft. The Firth of Forth Bridge (Figure 3.7) is made of a pair of cantilevered arms "sticking out" from two main towers. The beams are supported by diagonal steel tubes projecting from the top and bottom of the towers. These well-secured spans actually support the central span. This design makes the Firth of Forth Bridge one of the strongest—and most expensive—bridges ever built. The bridge was opened in 1890.

Figure 3.7. The Firth of Forth Bridge was completed in 1890 in the United Kingdom.

The Firth of Forth Bridge, which was thought to have been overdesigned, was the basic model for the Quebec Bridge. This bridge across the St. Lawrence River was the brainchild of the Quebec Bridge Company. In 1903, the Quebec Bridge Company gave the job of designing of the bridge to the Phoenix Bridge Company. The company also contracted a renowned bridge builder, Theodore Cooper from New York, to oversee the engineering design and construction.

The peculiarities of the site made the design of the bridge most difficult. Because the St. Lawrence was a shipping lane, the 2800-foot bridge was required to have a 1800-foot single span to allow the oceangoing vessels to pass. Further, the bridge was to be multifunctional and 67   ft. wide to accommodate two railway tracks, two streetcar tracks, and two roadways. The key to the cantilevered bridge design was the weight of the center span.

In late 1903, P.L. Szlapaka of the Phoenix Bridge Company laid out the initial drawings for the bridge. His design, which had the background of the overdesigned Firth of Forth Bridge, was approved with very few changes. The estimated weight of the span was calculated based on these initial drawings. In 1905, the working drawings were completed and the first steel girder was bolted into place. These working drawings took over 7 months for final approval. In the meantime, the work had begun. It was not until Cooper received the drawing that he noticed that the estimated weight of the span was off, on the low side, by almost 8 million pounds. Cooper had two choices: condemn the design and start over or take a risk that there would be no problem. Telling himself that the 8 million pounds was within engineering tolerances, Cooper let the work continue (Figure 3.8). After all, he wanted to be known as the designer of the greatest bridge in the world. Also factored into the decision was that the Prince of Wales (later to be King George V) was scheduled to open the bridge in 1908 and any delay in construction would upset the planning.

Figure 3.8. The Quebec Bridge under construction in 1907.

On June 15, 1907, an inspecting engineer noted that two girders of the anchor were misaligned by a quarter of an inch. Cooper called this a "not serious" problem. In the inspection report in August 1907, it was noted that the girders had moved out of alignment a bit more and "appeared bent." Although this condition was a bit more disconcerting, the work continued. On August 27, 1907, the warning bells finally went off when the inspection team noted that, over the weekend, the girders had shifted a "couple of inches" and were more obviously bent. At 5:32 p.m. on August 29, the girders trembled with a grinding noise and gave way. The bridge structure plunged over 150   ft., taking with it the lives of 75 workers (Figure 3.9).

Figure 3.9. The collapsed Quebec Bridge.

The members of the Royal Commission of Inquiry investigating the collapse wrote in their 1908 report, "A grave error was made in assuming the dead load for the calculations at too low a value... This error was of sufficient magnitude to have required the condemnation of the bridge, even if the details of the lower chords had been of sufficient strength." The lower chord members had a rectangular section of 5   ft., 7.5   in., not much smaller than the overdesigned Firth of Forth Bridge chord's circular intersection. The load at the center span was much greater than anticipated, and even though the lower chord section was comparable to the Firth of Forth Bridge chord, it was not sufficient due to the size of the middle span. Clearly, the scale of the center span should have been considered and the chord section enlarged. Extrapolation from other bridge data simply did not work in this case. The bridge finally was completed with the help from the British engineer who had worked on the Firth of Forth Bridge.

The significance of the scale factor also is reflected when we wonder why we do not have any giants. To be able to stand up, the bones of a 40-foot giant must support a load that would require bones of much bigger size or of a different material; larger bones would also interfere with limb movements.

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Design Examples of Steel and Steel-Concrete Composite Bridges

Ehab Ellobody , in Finite Element Analysis and Design of Steel and Steel-Concrete Composite Bridges, 2014

4.5.9 Design of Wind Bracings

Wind forces acting on the double-track railway bridge (see Figure 4.152) as well as any other lateral forces directly applied to the bridge are transmitted to the bearings by systems of wind bracing. For pony bridges, only the lower wind bracing carries wind forces on the moving train, wind forces on the main plate girder, and lateral shock (nosing force) applied to the tracks (see Figure 4.153). Wind forces applied to this bridge can be sufficiently estimated using the design rules specified in EC1 [3.2] as follows:

Figure 4.152. Design heights for the calculation of wind forces on the lower wind bracings.

Figure 4.153. Loads on the lower wind bracing.

$F_{\mathrm{w}}=\frac{1}{2}\rho v_{\mathrm{b}}^{2}C{A}_{\mathrm{ref},x}$

$v_{\mathrm{b}}=c_{\mathrm{dir}}\times c_{\mathrm{season}}\times v_{\mathrm{b},0}=1.0\times 1.0\times 26=26\mathrm{m}/\mathrm{s}$

$A_{\mathrm{ref},x}=4.9118\times 28=137.53{\mathrm{m}}^2$

$F_{\mathrm{w}}=\frac{1}{2}\times 1.25\times {26}^2\times 5.7\times 137.53=331,208\mathrm{N}=331.2\mathrm{kN}$

Considering the structural analysis for the lower wind bracing system shown in Figure 4.153, the critical design wind force in the diagonal bracing members can be calculated as follows:

$\mathrm{Distributed}\mathrm{wind}\mathrm{loads}\left(q_{\mathrm{WL}}\right)=522.6\times \left(5.5/7\right)/30=13.69\mathrm{kN}/\mathrm{m}$

$\mathrm{Factored}\mathrm{distributed}\mathrm{wind}\mathrm{loads}=q_{\mathrm{WL}}\times {\gamma }_{\mathrm{q}}=13.69\times 1.7=23.27\mathrm{kN}/\mathrm{m}$

$R_{\mathrm{A}}=100+20.86\times 13.5=381.6\mathrm{kN}$

$\alpha ={tan}^{-1}\left(4.5/4.5\right)=45°$

$F_{\mathrm{D}}=281.6/\left(2\times sin45\right)=199.1\mathrm{kN}$

The cross section of the bracing member (see Figure 4.154) can be determined as follows:

Figure 4.154. Upper wind bracing cross section s-s.

$l_{\mathrm{b}x}=6360\mathrm{mm},l_{\mathrm{b}y}=1.2\times 6360=7632\mathrm{mm}$

Choose two angles back-to-back 150   ×   150   ×   15, with 10   mm gusset plate between them:

$A=2\times 43.2=86.4{\mathrm{cm}}^2,i_x=4.59\mathrm{cm},e=4.26\mathrm{cm},i_y=\sqrt{{4.59}^2+{\left(4.26+1/2\right)}^2}=6.61\mathrm{cm}$

$\varepsilon =\sqrt{\frac{235}{275}}=0.924$

$\overline{\lambda }=\frac{L_{\mathrm{cr}}}{i}\frac{1}{{\lambda }_1}$

${\lambda }_1=93.9\times 0.924=86.7636$

$\overline{\lambda }=\frac{6360}{45.9}\frac{1}{86.7636}=1.597$

The axial compressive force in the diagonal bracing member (N _Ed  =   298.7   kN):

$\frac{N_{\mathrm{Ed}}}{N_{\mathrm{b},\mathrm{Rd}}}\le 1.0$

where $N_{\mathrm{b},\mathrm{Rd}}=\frac{\mathit{\chi A}{f}_{\mathrm{y}}}{{\gamma }_{\mathrm{M}1}}$

$A=2\times 43.2=86.4{\mathrm{cm}}^2$

$\chi =\frac{1}{\Phi +\sqrt{{\Phi }^2-{\overline{\lambda }}^2}}\mathrm{but}\chi \le 1.0$

$\Phi =0.5\left[1+\alpha \left(\overline{\lambda }-0.2\right)+{\overline{\lambda }}^2\right]$

$\Phi =0.5\left[1+0.34\left(1.597-0.2\right)+{1.597}^2\right]=2.013$

$\chi =\frac{1}{2.013+\sqrt{{2.013}^2-{1.597}^2}}=0.309\mathrm{but}\chi \le 1.0$

Then, $N_{\mathrm{b},\mathrm{Rd}}=\frac{0.309\times 86.4\times 100\times 275}{1.1}=667,440\mathrm{N}$

$N_{\mathrm{b},\mathrm{Rd}}=667.4\mathrm{kN}>N_{\mathrm{Ed}}=199.1\mathrm{kN}\left(\mathrm{Then}\mathrm{O}.\mathrm{K}.\right)$

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Arch bridges

F. Schanack , O.R. Ramos , in Innovative Bridge Design Handbook, 2016

3.3 Review of recent innovative tied arch bridges

Tied arches, together with trusses, have traditionally been used for railway bridges when long spans are required. In HSR lines, the allowed bridge deflection is very limited. Therefore, tied arches for HSR bridges incorporate innovative design features to increase the stiffness of the structure. In the Mornas Viaduct (erected in France in 1999 and featuring a span of 121.4  m) and the Xinkaihe Bridge (built in China in 2005, with a span of 140   m), this was achieved by double arches. In other bridges, such as the Rego Das Lamas viaduct (built in Spain in 2012, with a span of 80   m), the Yichang Railway Bridge (built in China in 2008, with a span of 275   m) or the Donzére-Mondragon Viaduct (built in France in 1999, with a span of 110   m), the deck is a box girder. The network arch itself is very stiff, but for HSR traffic, the arches are less slender than usual. For example, in the Huihe River Bridge of the Beijing-Shanghai railway line (in China in 2011, with a span of 96   m) and in the East Lake Bridge of the Wuhan-Guangzhou railway line (in China in 2010, with a span of 112   m) the arches are made of two CFSTs each (Hu et al., 2014).

From a global point of view, CFST arches can still be considered innovative, although the first CFST arch was built in 1936 and there are currently more than 150 CFST tied arches in China. This technique combines the easy erectability of steel arches with the benefit of using concrete (poured after arch assembly) for handling compressive forces. Outside of China, very few CFST tied arch bridges have been built; one example is the Ravenna Viaduct in Nebraska, built in 2005, with a 53-m span).

In the first half of the 20th century, concrete has been used very frequently for arch bridges, but was then replaced by steel due to high labor costs for the arch formwork. There are very few recent bridges with concrete arches, like the Norridgewock Bridge in Maine (built in 2011, with a span of 91.44   m), the Depot Street Bridge in Oregon (built in 2006, with a span of 100   m) and the Third Millennium Bridge in Zaragoza, Spain (built in 2008, with a span of 216   m). The concrete arches of theses bridges were cast in place on scaffolding. A very innovative technology was used for the 7th Street Bridge in Ft. Worth, TX (built in 2013, with a span of 49.68   m). It has six spans and a total of 12 identical concrete network arches, which were precast off site (Figure 13.29).

Figure 13.29. Precasting and rotation of the arches, 7th Street Bridge, Fort Worth, TX.

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Repair, Strengthening, and Replacement

Weiwei Lin , Teruhiko Yoda , in Bridge Engineering, 2017

14.5.1 Strengthening Method Description

The proposed method aims to be used for strengthening short-span steel railway bridge superstructures and longitudinal-lateral beam connection in plate girder bridges. The philosophy of the proposed method is to change the steel section to be the composite section by integrating with new materials, which in turn reduces stress ranges and extend the residual fatigue life of the aged structures. New materials including rubber-latex, rapid hardening concrete, reinforcement, and glass fiber-reinforced polymer (GFRP) plates are used. In the first step, the old structural steel is cleaned and then rubber-latex mortar is sprayed on the surface of the structural steel for protecting the structural steel from corrosion, increasing the bond strength on the steel-concrete interface, and reducing the noise in the service stage. Concrete and mortar, including styrene-butadiene rubber- latex, show various abilities especially in adhesion bonding, waterproofing, and shock absorption and abrasion resistance. In the second step, reinforcing bars and GFRP plates should be installed. GFRP plates are very easy to carry due to their light-weight. In this method, the GFRP plates are used as formworks for concrete casting. Then after that, the light-weight rapid hardening concrete is poured to finish the strengthening. For the maintenance or repairing of the railway bridge, rapid construction and light weight are the critical points. In this strengthening method, the concrete casting can be scheduled in the night time and will not affect the public traffic. The real effects of the present methods were confirmed by following application examples.

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Used Double Track Bombardier in Ny State

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